Design Of Spillways

 DESIGN OF SPILLWAYS 

1.0 GENERAL 

 

Spillway is a safety valve provided in the dam to dispose of surplus flood waters from a reservoir after it has been filled to its maximum capacity i.e. Full Reservoir Level. 

 

The importance of safe spillways needs no over-emphasis as many failures of dams have been caused by improper design of spillways or spillways of insufficient capacity especially in case of earth and rockfill dams which are susceptible to breaching, if overtopped.  Concrete/Masonry dams can withstand moderate overtopping but this should be avoided. 

 

Further, the spillway must be hydraulically and structurally adequate and must be so located that the overflowing discharges do not erode or undermine the downstream toe of the dam. 

 

2.0 SELECTION OF DESIGN FLOOD  

 

The spillway design flood is generally determined by transposing great storms which have been known to occur in the region over the drainage area. The resulting flood hydrographs are then determined by rational methods. In determining the discharge capacity consideration should be given to all possible contingencies, e.g. one or more gates being inoperative.  

 

IS : 11223 – 1985 provides guidelines for fixing spillway capacity.  Inflow design flood for the safety of the dam is guided by the following criterion: 

 

The dams may be classified according to size by using the hydraulic head and the gross storage behind the dam as given below.  The overall size classification for the dam would be the greater of that indicated by either of the following two parameters: 

 

Classification  

  Gross Storage Hydraulic Head 

Small          0.5 to 10 million m³ 7.5 to 12 m  

Intermediate        10 to 60 million m³ 12 to 30m 

Large          Above 60 million m³  Above 30 m 

 

The inflow design flood for safety of the dam would be as follows: 

 

Size of dam Inflow Design Flood for safety of dam 

 

Small  100 years flood 

Intermediate  Standard Project Flood (SPF) 

Large  Probable Maximum Flood (PMF) 

 

Floods of larger or smaller magnitude may be used if the hazard involved in the eventuality of a failure is particularly high or low.  The relevant parameters to be considered in judging the hazard in addition to the size would be: 

 

i) Distance and location of human habitations on the downstream after considering the likely future developments 

ii) Maximum hydraulic capacity of the downstream channel 

 

For more important projects dam break studies are done as an aid to the judgement in deciding whether PMF needs to be used.  Where the studies or judgement indicate an imminent danger to present or future human settlements, the PMF should be used as design flood. 

 

2.1 Standard Project Flood (SPF) 

 

It is the flood that may be expected from the most severe combination of hydrological and meteorological factors that are considered reasonably characteristic of the region and is computed by using the Standard Project Storm (SPS).  While transposition of storms from outside the basin is permissible, very rare storms which are not characteristic of the region concerned are excluded in arriving at the SPS rainfall of the basin. 

 

2.2 Probable Maximum Flood (PMF) 

 

It is the flood that may be expected from the most severe combination of critical meteorological and hydrological condition that are reasonably possible in the region and is computed by using the Probable Maximum Storm (PMS) which is an estimate of the physical upper limit to maximum precipitation for the basin.  This is obtained from the transposition studies of the storms that have occurred over the region and maximizing them for the most critical atmospheric conditions. 

 

3.0 FLOOD ROUTING 

 

The process of computing the reservoir storage volumes and outflow rates corresponding to a particular hydrograph of inflow is known as flood routing. It is used for arriving at the MWL for the project.  The relationship governing the computation is essentially simple – over any interval of time the volume of inflow must equal the volume of outflow plus the change in storage during the period. If the reservoir is rising, there will be increase in storage and change in storage will be positive, if the reservoir is falling, there will be decrease in storage and the change in storage will be negative. 

 

For an interval of time Δt, the relationship can be expressed by the following expression: 

 

  I.  t

Where,

  I = Average rate of inflow during time equal to Δt 

  O = Average rate of outflow during time equal to Δt 

ΔS = Storage accumulated during time equal to Δt 

 

The following three curves are required for carrying out the computations: 

 

a) The inflow flood hydrograph 

b) The reservoir capacity curve  

c) The rating curve showing the total rate of outflow through outlets and over the spillway against various reservoir elevations 

 

Flood routing in gated spillways is generally carried out assuming the flood to impinge at FRL assuming inflow equal outflow to at that level.  For ungated spillways this would correspond to the spillway crest or a little above this.   

 

The methods generally adopted for flood routing studies are:  

 

i) Trial and Error Method 

ii) Modified Puls Method 

 

3.1 Trial and Error Method 

 

This method arranges the basic routing equation as follows: 

 

I1

 

  2 2

The procedure involves assuming a particular level in the reservoir at the time interval Δt, and computing the values on the right side of the above equation. The computed value on the right side of the equation, corresponding to the assumed reservoir level, is compared with the known value on the left side of the equation. If the two values tally, then the assumed reservoir level at the end of the time interval is OK; otherwise a new reservoir level is assumed and the process is repeated till the required matching is obtained. 

 

This method gives quite reliable results, provided the chosen time interval is sufficiently small, so that the mean of the outflow rates at the start and the end of the interval may be taken as the average throughout the interval. 

 

3.2 Modified Puls Method 

 

This method arranges the basic routing equation as below so that the knowns are placed on the left side and unknowns are placed on the right side of the equation. 

 

Since this equation contains two unknowns it cannot be solved unless a second independent function is available. In the modified Puls method, a storage-indication curve viz. outflow O versus the quantity (S/Δt+O/2) is constructed for the purpose. 

 

In the above equation, it may be noted that subtracting O2 from (S2/Δt+O2/2) gives (S2/ΔtO2/2). This expression is identical to (S1/Δt-O1/2) on the left side of the equation except for the subscripts. Since the subscript 1 denotes values at the beginning of a time increment and subscript 2 denotes values at the end of a time increment, it is apparent that (S2/Δt-O2/2) at the end of one time increment is numerically equal to (S1/Δt-O1/2) for the beginning of the succeeding time increment. 

 

The detailed routing procedure is as follows: 

 

i) Compute the numerical value of left side of the equation for given values of I1, I2, S1 and O1 for the first time increment. 

 

ii) With this numerical value, which equals (S2/Δt+O2/2), refer storage-indication curve and read outflow O2 corresponding to this computed value of (S2/Δt+O2/2). The O2 thus read is the instantaneous outflow at the end of the first time increment. 

 

iii) Subtract this value of O2 from (S2/Δt+O2/2), which gives the value for (S2/ΔtO2/2). The value of (S1/Δt-O1/2) for the second time increment is equal to (S2/ΔtO2/2) for the first time increment. Consequently the left side of the equation can be computed for the second time increment and the entire procedure is repeated. 

      TYPES OF SPILLWAYS 

Spillways can be classified as controlled or uncontrolled depending upon whether they are gated or ungated.  Further they are also classified based on other prominent features such as control structure, discharge channel or some such other components.   

The common types of spillways used are: 

i) Overfall or Ogee  ii) Orifice or sluice iii) Chute or trough iv) Side channel  

v) Tunnel/Shaft or Morning Glory  vi) Siphon  

4.1 Overfall or Ogee Spillway 

 

The overfall type is by far the most common and is adapted to masonry dams that have sufficient crest length to provide the desired capacity. 

 

This type comprises a control weir which is ogee or S-shaped.  The ogee shape conforms to the profile of aerated lower nappe from a sharp crested weir.  The upper curve at the crest may be made either broader or sharper than the nappe.  A broader curve will support the sheet and positive hydrostatic pressure will occur along the contact surface.  The support sheet thus creates a backwater effect and reduces the coefficient of discharge.  The sharper crest on the other hand creates negative pressures, increases the effective head and thereby the discharge. 

 

These spillways are generally provided in Masonry/Concrete dams and also in composite dams as central spillways located in the main river course.  Examples are Bhakra dam, Rihand dam, Sriram Sagar dam, Nagarjunasagar dam, Jawahar Sagar dam, Tenughat dam, Srisailam dam, Tawa dam, Ukai dam etc. 

 

A typical ogee spillway is shown in figure below: 

 

  

4.2 Orifice Spillway 

 

Low crested spillways with either breast wall or sluice type arrangements are now increasingly being provided for flushing out the silt and controlling the silt entry in the power intake which is kept above the spillway crest.  These spillways are called orifice or sluice spillways (See figures below). 

 

  

  

The orifice spillways have the advantage of having high discharging capacity due to the high water head. At sites where only a limited area and relatively short length of suitable foundation material are available for the spillway structure, the orifice spillway offers the most economic means of passing the design flood. However, the orifice spillways result in high flow concentration, which increases the size and cost of energy dissipation work below. 

 

Orifice spillways are being provided in many of our diversion dams recently in rivers carrying heavy silt load.  The power intake is kept above the spillway crest and as close to the spillway as possible.  This kind of spillway arrangements thus performs the dual function of passing the flood and managing the sediment in the reservoir. 

 

Examples of spillways with breast wall type arrangements are in Ranganadi H.E. Project (Arunachal Pradesh), Chamera H.E. Project Stage-I (Himachal Pradesh), Rangit H.E. Project Stage-III (Sikkim) etc. and that of sluice spillways are in Tala H.E. Project (Bhutan), Nathpa Jhakri H.E. Project (Himachal Pradesh), Myntdu H.E. Project Stage-I (Meghalaya) etc. 

 

4.3 Chute Spillway 

 

A spillway where discharge is conveyed from the reservoir to the downstream river through an open channel or chute along a dam abutment or through a saddle is called a chute or trough spillway.  The chute is the commonest type of water conductor used for conveying flow between control structures and energy dissipators. Chute can be formed on the downstream face of gravity dams, cut into rock abutments and either concretelined or left unlined and built as free-standing structures on foundations of rock or soil.  

 

These are mostly used with earth/rockfill dams and have the following main advantages: 

 

 

i) Simplicity of design 

ii) Adaptability to all types of foundations and 

iii) Overall economy by using large amount of spillway excavation in dam construction 

 

Examples of chute spillways are Beas Dam, Ram Ganga Dam, Kolar Dam, Tehri Dam etc. 

 

A typical chute spillway is shown in figure below: 

 

  

 

4.4 Side Channel Spillway 

 

The distinctive feature of side channel spillway which distinguishes it from chute spillway is that whereas in the chute spillway the water flows at right angle to the axis of the dam, in the side channel spillway, the flow is initially in a channel parallel to the axis of the dam and thereafter it flows in a discharge channel at approximately right angle to the dam axis. 

 

This type of spillway is suited to narrow canyons with steep sides which rise to a considerable height above the dam. This type of spillway is also provided at sites where the overfall type is ruled out for some reason and where saddle of sufficient width is not available to accommodate a trough (chute) type spillway. It is assumed that all the energy of the overfalling water is dissipated in turbulence in the side channel. Example of side channel spillway is Pancheshwar Project. 

 

 

A typical layout of a side channel spillway is illustrated in figure below: 

 

  

4.5 Tunnel/Shaft or Morning Glory Spillway 

 

In this type of spillway water enters over the lip of a horizontal circular crest and drops through a vertical or sloping shaft and then flows downstream through a horizontal conduit or tunnel.  The spillway is suitable to dam sites in narrow canyons where room for a spillway restricted.  

 

In some instances advantage of the existing diversion tunnel has been taken for conversion into tunnel spillway. A disadvantage of this type is that the discharge beyond a certain point increases only slightly with increased depth of overflow and therefore does not give much factor of safety against underestimation flood discharge as compared to the other types. 

 

Examples of Tunnel/Shaft spillways are Tehri Dam, Itaipau Dam etc.  A typical Tunnel/Shaft Spillway is shown in figure below: 

 

  

 

4.6 Siphon Spillway 

 

Siphon spillways are based on the principle of siphonic action in an inverted bent pipe.  If such pipe is once filled with water, it will continue to flow so long as the liquid surface is higher than the lower leg of the pipe unless of course, the upper leg gets exposed earlier.  

 

Siphon spillways are often superior to other forms where the available space is limited and the discharge is not extremely large. They are also useful in providing automatic surface-level regulation within narrow limits. The siphon spillways prime rapidly and bring into action their full capacity. Therefore, they are especially useful at the power house end of long power channels with limited forebay capacity where a considerable discharge capacity is necessary within a very short time in order to avoid overflow of the channel banks.   

 

However, siphon spillways are not very common mainly because of: 

 

i) Possibility of clogging of the siphon passage way and siphon breaker vents with debris, leaves etc. 

ii) The occurrence of sudden surges and stoppages of outflow as a result of the erratic make and break action of the siphon, thus causing fluctuations in the downstream river stage. 

iii) The release of outflows in excess of reservoir inflows whenever the siphon operates, if a single siphon is used.  Closer regulation which will more nearly balance outflow and inflow can be obtained by providing a series of smaller siphons with their siphon breaker vents set to prime at gradually increasing reservoir heads. 

iv) Vibration disturbances which are more pronounced than in other types of spillways. 

 

A siphon spillway through a dam is shown in figure below: 

                         

 

 

4.0 HYDRAULIC DESIGN OF OVERFALL OGEE SPILLWAYS (Refer IS: 

6934) 

 

Overfall ogee spillway has its overflow profile conforming, as nearly as possible, to the profile of the lower nappe of a ventilated jet of water over a sharp crested weir. These spillways are classified as high and low depending on whether the ratio of height of the spillway crest measured from the river bed to the design head is greater than or equal or less than 1.33 respectively.  In the case of high overflow spillways the velocity of approach head may be considered negligible. 

 

5.1 Shape of Ogee Profile 

 

i)  Spillways with vertical upstream face 

 

Upstream Quadrant  

The upstream quadrant of the crest may conform to the ellipse:  

 

 

The magnitude of A1 and B1 are determined from the graph P/Hd vs A1/Hd and B1/Hd respectively in fig.2 of IS:6934,  where, 

 

P   = Height of crest from the river bed 

Hd = Design Head 

 

 

Downstream Quadrant   

The downstream profile of the crest may conform to the equation: 

 

 X 21.85K 2 .Hd 0.85 .Y2

The magnitude of K2 is determined from the graph P/Hd vs K2 in fig.2  of IS:6934. 

 

ii)  Spillways with sloping upstream face 

 

In the case of sloping upstream face, the desired inclination of the face is fitted tangentially to the elliptical profile described under (i) above, with the appropriate tangent point worked out from the equation.  The profile of the downstream quadrant remains unchanged. 

 

  

  

Figure 2 – IS:6934 

 

 

 

 

iii) Spillways with crest offsets and risers  

 

Whenever structural requirements permit, removal of some mass from the upstream face leading to offsets and risers as shown in fig.1 of IS:6934 results in economy.  The ratio of risers M to the design head Hd i.e. M/Hd should be at least 0.6 or larger, for the flow condition to be stable.  The shapes of u/s and d/s quadrants defined for spillways with vertical upstream face are also applicable to overhanging crests, for the ratio M/Hd > 0.6. 

  

5.2 Discharge Computations 

 

The discharge over the spillway is generally computed by the equation 

 

  Q 2 gC.L.H 3/ 2

where, 

  C = Coefficient of Discharge 

  L = Effective length of crest 

  H = Head over crest 

 

i) Effective Length of Overflow Crest  

 

The net length of overflow crest is reduced due to contraction caused by abutment and crest piers.  The effective length L of the crest may be calculated as follows: 

 

    2H(N.Kp Ka)

where,

  L’ = Overall length of the crest excluding piers 

  H = Head over crest 

  N = Number of piers 

  Kp = End contraction coefficient of piers 

  Ka = End contraction coefficient of abutment 

 

The pier contraction coefficient, Kp is affected by the shape and location of the pier nose, thickness of pier, the actual head in relation to the design head and the approach velocity. The average pier contraction coefficients may be taken as follows: 

 

Type Kp 

 

Square-nosed piers with rounded corners of radius about 0.1 times pier thickness 

  0.02 

Round-nosed piers 0.01 

 

Pointed-nosed piers 0.0 

 

The abutment contraction coefficient, Ka is affected by the shape of the abutment, the angle between the upstream approach wall and the axis of flow, the actual head in relation to the design head and the approach velocity. The average abutment coefficient may be taken as follows: 

  

Type Ka 

 

Square abutments with head wall at 90o  to direction of flow 0.2 

Rounded abutments with head wall at 90o  to direction of flow, when 0.5Hd > R > 

0.15Hd 

  0.1 

 

Rounded abutments where R > 0.5Hd and head wall not more than 45o  to direction of flow 0.0 

 

ii) Coefficient of Discharge 

 

The value of coefficient of discharge depends on the following: 

 

a) Shape of the crest 

b) Depth of overflow in relation to design head 

c) Depth of approach 

d) Extent of submergence due to tail water 

e) Inclination of the upstream face 

 

Fig. 3 of IS:6934 gives the coefficient of discharge C for the design head as a function of approach depth and inclination of upstream face of the spillway.  These curves can be used for preliminary design purposes. 

  

Fig. 4 of IS:6934 gives the variation of coefficient of discharge as a function of ratio of the actual head to the design head (i.e. H/Hd).  This curve can be used to estimate C for heads other than design head. 

 

  

 

The coefficient of discharge is reduced due to submergence by the tail water.  The position of the downstream apron relative to the crest level also has an effect on the discharge coefficient.  Fig. 5A and 5B of IS:6934 give the variation of C with the above parameters. 

  

iii) Design Head 

When the ogee crest is formed to a shape differing from the ideal shape or when the crest has been shaped for a head larger or smaller than the one under consideration, the coefficient of discharge will differ.   

A design head grater than the actual head will push the crest surface into the theoretical nappe and result in greater pressure along the curved surfaces and in lower discharge capacities. Conversely, a design head lower than the actual head pulls the crest surface below the theoretical nappe, resulting in sub-atmospheric pressures over some portion of the crest curve. At the same time the discharge capacity of such a crest curve is increased.  

Excessive sub-atmospheric pressures can result in pulsating, inefficient spillway operation, and possibly damage to the structure as a result of cavitations. A certain amount of sub-atmospheric pressure can be attained without undesirable effects. Figure provides a guide for determining the minimum pressures on the crest for various ratios of design head and actual head on the crest. 

  

Designing the crest shape to fit the nappe for a head less than maximum head expected often results in economies in construction. The resulting increase in unit discharge may make possible a shortening of the crest length, or a reduction in freeboard allowance for reservoir surcharge under extreme flood conditions.  

Because the occurrence of design floods is usually so infrequent, the spillway crests are fitted to the lower nappe of a head which is 75% of that resulting from the actual discharge capacity. Tests have shown that the sub-atmospheric pressures on a nappeshaped crest do not exceed about half of the design head when the design head is not less that about 75% of the maximum head. An approximate diagram of the sub-atmospheric pressures, as determined from model tests, is shown by figure. The design head is normally kept as 80% to 90% of the maximum head corresponding to MWL. 

  

The minimum crest pressure must be greater than cavitation pressure. It is suggested that the minimum pressure allowable for design purposes be 20ft of water below sea level atmospheric pressure and that the altitude of the project site be taken into account in making the calculation. For example, assume a site where the atmospheric pressure if 5ft of water less than sea level pressure, and in which the maximum head contemplated is 60ft; then, only 15 additional feet of sub-atmospheric pressure is allowable.  

Expansion Joint

 Typical Components of an Expansion Joint

In a typical expansion joint, it normally contains the following components: 

joint sealant, 

joint filler, 

dowel bar, 

PVC dowel sleeve, 

bond breaker tape and cradle bent. 

Joint sealant: it seals the joint width and prevents water and dirt from entering the joint and causing dowel bar corrosion and unexpected joint stress resulting from restrained movement. 

Joint filler: it is compressible so that the joint can expand freely without constraint. Someone may doubt that even without its presence, the joint can still expand freely. In fact, its presence is necessary because it serves the purpose of space occupation such that even if dirt and rubbish are intruded in the joint, there is no space left for their accommodation. 

Dowel bar: This is a major component of the joint. It serves to guide the direction of movement of concrete expansion. Therefore, incorrect direction of placement of dowel bar will induce stresses in the joint during thermal expansion. On the other hand, it links the two adjacent structures by transferring loads across the joints. 

PVC dowel sleeve: It serves to facilitate the movement of dowel bar. On one side of the joint, the dowel bar is encased in concrete. On the other side, however, the PVC dowel sleeve is bonded directly to concrete so that movement of dowel bar can take place. One may notice that the detailing of normal expansion joints in Highways Standard Drawing is in such a way that part of PVC dowel sleeve is also extended to the other part of the joint where the dowel bar is directly adhered to concrete. In this case, it appears that this arrangement prevents the movement of joint. If this is the case, why should designers purposely put up such arrangement? In fact, the rationale behind this is to avoid water from getting into contact with dowel bar in case the joint sealant fails. As PVC is a flexible material, it only minutely hinders the movement of joint only under this design. 

Bond breaker tape: As the majority of joint sealant is applied in liquid form during construction, the bond breaker tape helps to prevent flowing of sealant liquid inside the joint . 

Freeboard In Dams

 Freeboard

Free Board is the vertical distance between the top of the dam and the still water level. Freeboard is computed from the following two considerations:

Wave height considerations

It is equal to wind set up plus 1 1/3 times the wave height above FRL or above MWL (corresponding to design flood) whichever gives higher dam top level. A minimum freeboard of 1m above MWL corresponding to design flood shall be available. If design flood is not equal to PMF then the top of dam should be at least equal to MWL corresponding to PMF. At least 1m high solid parapet is to be provided, not withstanding the above requirements.

Wind velocity generally assumed as below in absence of meteorological data:

For FRL condition - 120 km/hr

For MWL condition - 80 km/hr

T. Saville’s method as given in IS:6512-1984 is used for calculating the wave height/freeboard.

Operation considerations

IS:11223 specifies the following:

The freeboard as specified in IS: 6512 shall be available at FRL and MWL corresponding to all bays operative condition. For gated spillways a contingency of 10% of gates (min. one gate) being inoperative is considered as an emergency. A reduced freeboard may be acceptable under the emergency condition. The dam shall not be allowed to overtop in any case. 

Stability Analysis of Concrete Gravity Dams

 

4.0 Design Criteria - Stability Analysis 

4.1 Requirements for Stability

The following are the basic requirements of stability for a gravity dam:

 

i) Dam shall be safe against sliding at any section in the dam/dam foundation interface/within the foundation.

ii) Safe unit stresses in concrete/masonry shall not be exceeded.

iii) Dam shall be safe against overturning

4.2 Basic Assumptions 

For stability analysis the following assumptions are made:

i) That the dam is composed of individual transverse vertical elements each of which carries its load to the foundation without transfer of load from or to adjacent elements.

ii) That the vertical stress varies linearly from upstream face to downstream face on any horizontal section.

4.3 Load Combinations

The following loading combinations have been prescribed by IS:6512 for stability analysis:

Combination A (Construction condition) Dam completed but no water in reservoir and no tail water

Combination B (Normal operating condition) Reservoir at full reservoir elevation, normal dry weather tail water, normal uplift and silt.

Combination C (Flood Discharge condition) Reservoir at maximum flood pool elevation, all gates open, tail water at flood elevation, normal uplift and silt

Combination D Combination A with earthquake

Combination E Combination B with earthquake 

Combination F Combination C with extreme uplift (Drains inoperative)

Combination G Combination E with extreme uplift (Drains inoperative)

4.4 Forces acting on a gravity dam

The various external forces considered to be acting on a gravity dam are:

1. Dead loads

2. Reservoir and Tail water loads

3. Uplift pressures

4. Earthquake forces

5. Silt pressures

6. Ice pressure

7. Wave pressure

8. Thermal loads, if applicable

Dead Loads

Self weight of dam and weight of appurtenant works such as, spillway piers, gates, hoists, spillway bridge etc. are considered for computing the dead loads. The unit weights adopted in preliminary designs are 2.3 t/m3 for masonry and 2.4 t/m3 for concrete.

Reservoir & Tail Water Loads

The load due to reservoir water is calculated using hydrostatic triangular pressure distribution and taking unit weight of water as 1 t/m3. The weight of flowing water over spillway is neglected. The load due to tail water is calculated by taking tail water pressure corresponding to tail water elevation in case of NOF sections and for a reduced value in case of OF sections depending on the E.D.A.

Silt Pressures

The deposited silt may be taken as equivalent to a fluid exerting a force with a unit weight 0.36 t/m3 in horizontal direction and 0.925 t/m3 in vertical direction. Thus the horizontal silt and water pressure is determined as if silt and water have a horizontal unit weight of 1.36 t/m3 and vertical silt and water pressure is determined as if silt and water have a vertical unit weight of 1.925 t/m3.

Uplift Pressures

Water seeping through the pores, cracks and fissures of the foundation material, and water seeping through the body of the dam exert an uplift pressure on the base of the dam. It is assumed to act over 100% of the area of base and assumed to vary linearly from upstream to downstream corresponding to water heads. However, in case drainage galleries are provided, there is a relief of uplift pressure at the line of drain equal to two-thirds the difference of the hydrostatic heads at upstream and downstream. It is assumed that uplift pressures are not affected by earthquakes.

Earthquake forces (Ref: IS 1893 – 1984)

Earthquake forces are determined as per IS:1893. Design seismic coefficients in horizontal and vertical direction are worked out as per the above code based on the location of the project on the seismic map of India.  

Design horizontal seismic coefficient ( h)

1. By Seismic Coefficient Method (for dams upto 100 m high)

 h  =   I  o

where,

 h  = Design Horizontal Seismic Coefficient

 o = Basic Horizontal Seismic Coefficient (from Table 2, IS:1893)

I = Importance factor of the structure (3.0 for dams)

 = Coefficient depending upon soil foundation system

(1.0 for dams)

2. By Response Spectrum Method (For dams higher than 100 m)

 h  =  I FoSa/g

where,

Fo = Seismic Zone factor for average acceleration spectra (from 

Table 2, IS:1893)

Sa/g = Average Acceleration coefficient read from Fig. 2, IS:1893 for 

appropriate natural period and damping.

Design vertical seismic coefficient 

The design vertical seismic coefficient is taken as half the design horizontal seismic coefficient.

Inertia forces on the dam

A triangular distribution of acceleration is prescribed for determining inertia forces on the dam.  For horizontal inertia forces 1.5 times the design horizontal seismic coefficient is assumed at the top of the dam varying to zero at the base.  For vertical inertia forces also 1.5 times the design vertical seismic coefficient is assumed at the top of the dam varying to zero at the base.  

Hydrodynamic Pressure on the dam

The basic work in this regard has been done by Westergaard.  Subsequently Zanger in 1952 presented formulas for computing hydrodynamic pressures exerted on vertical and sloping faces by horizontal earthquake effects.  Based on Zanger’s work, IS:1893 gives the procedure for calculating hydrodynamic pressure on the dam.

Effects of Horizontal Earthquake Acceleration

Due to horizontal acceleration of the foundation and dam there is an instantaneous hydrodynamic pressure (or suction) exerted against the dam in addition to hydrostatic forces.  The direction of hydrodynamic force is opposite to the direction of earthquake acceleration. Based on the assumption that water is incompressible, the hydrodynamic pressure at depth y below the reservoir surface shall be determined as follows :

          p = Cs h wh

where,

 p    =  hydrodynamic pressure in kg/m² at depth y,

Cs  =  coefficient which varies with shape and depth 

 h  =   design horizontal seismic coefficient 

w    =   unit weight of water in kg/m³, and

h    =   depth of reservoir in m.

The approximate values of Cs for dams with vertical or constant upstream slopes may be obtained as follows :

 

where,

Cm  =  maximum value of Cs obtained from Fig.10, IS:1893

y      =  depth below surface in m, and

h      =  depth of reservoir in m

 

Fig. 1 : Maximum Values of Pressure Coefficient (Cm)

for Constant Sloping Faces

The approximate values of total horizontal shear and moment about the center of gravity of a section due to hydrodynamic pressure are given by the relations :

           Vh   =  0.726 py

          Mh   =   0.299 py²

where

          Vh   =  hydroldynamic shear in kg/m at any depth, and

          Mh   = moment in kg.m/m due to hydrodynamic force at any depth y.    

Inertia forces on the dam

1. Seismic coefficient method (For dams upto 100 m high)

A triangular distribution of acceleration is prescribed for determining inertia forces on the dam.  For horizontal inertia forces 1.5 times the design horizontal seismic coefficient is assumed at the top of the dam varying to zero at the base.  For vertical inertia forces also 1.5 times the design vertical seismic coefficient is assumed at the top of the dam varying to zero at the base.  The design vertical seismic coefficient is taken as half the design horizontal seismic coefficient.

2. Response Spectrum Method (For dams more than 100 m high)

The fundamental period of vibration is calculated as under :

T = 5.55 H2/B (Wm/g/Es)0.5

where,

T = Fundamental period of vibration of the dam in secs.

H = Height of the dam in meters

B = Base width of the dam in meters

Wm = Unit weight of material of dam in kg/m

g = Acceleration due to gravity in m/sec

Es = Modulus of Elasticity of material in kg/m

Damping used for concrete dams = 5%

The design horizontal seismic coefficient is calculated using the above time period and for a damping of 5% from the average acceleration spectra given in IS:1893.

The basic shear and moment due to the horizontal inertia forces is obtained by the formulae given below:

Base shear = VB = 0.6 W.  h  

Base Moment = MB = 0.9 W.hCG  h  

where,

W = Self weight of dam in kg

hCG = Height of C.G. of dam above base in meters

 h  = Design Horizontal Seismic Coefficient

The vertical inertia forces are calculated using the same distribution as outlined in seismic coefficient method but using the seismic coefficient as calculated above.

5.0 Check for permissible stresses

 Check for Compressive Stresses

1. Concrete

• Strength of concrete after 1 year should be 4 times the maximum computed stress in the dam or 14 N/mm whichever is more.

• Allowable working stress in any part of the structure shall not exceed 7 N/mm.

2. Masonry

• Strength of masonry after 1 year should be 5 times the maximum computed strength in the dam or 12.5 N/mm which is more.

• Compressive strength of masonry can be determined by compressing to failure 75 cm cubes (or 45 cm x 90 cm cylinders) cored out of the structures.

Check for Tensile Stresses

Nominal tensile stresses permitted in concrete/masonry gravity dams (as per is: 6512)

Load Combination Permissible Tensile Stress

      Concrete dams                      Masonry dams

A Small Tension Small Tension

B No Tension No Tension

C 0.01 fc 0.005 fc

D Small Tension Small Tension

E 0.02 fc 0.01 fc

F 0.02 fc 0.01 fc

G 0.04 fc 0.02 fc

where,   fc = Cube Compressive Strength of Concrete/Masonry

6.0 Check for Sliding

The dam should be safe against sliding across any plane/combination of planes passing through:

- The body of the dam

- Dam foundation interface

- Foundation

The partial factors of safety against sliding as per IS:6512 are given below:

Loading

Condition F Fc

For dams and the Contact plane with Foundation For foundation

Thoroughly investigated Others

A,B,C 1.5 3.6 4.0 4.5

D,E 1.2 2.4 2.7 3.0

F,G 1.0 1.2 1.35 1.5

The factor of safety against sliding shall be computed from the following equation and it shall not be less than 1.0.

(W – U) tan    +  c.A

F =          F               Fc

                   P

Where,

F = factor of safety against sliding

W = total mass of the dam

U = total uplift force

tan  = coefficient of internal friction of the material

c = cohesion of the material at the plane considered

A = area under consideration for cohesion

F = partial factor of safety in respect of friction

Fc = partial factor of safety in respect of cohesion, and

P = total horizontal force

Design Of Concrete Gravity Dams - Introduction

 INTRODUCTION TO DESIGN OF CONCRETE GRAVITY DAMS

Introduction

A dam is an obstruction or a barrier built across a stream or a river for accumulation of water on its upstream side which is used for different purposes. Dams are constructed for deriving various benefits like irrigation, hydropower generation, flood control, domestic/industrial water supply, recreation etc. 

Dams can be classified based on various criteria. As per water resources planning the dams may be classified as storage dams, diversion dams and detention dams. As per hydraulic flow conditions the dams may be classified as overflow dams (spillways) and non-overflow dams. As per materials used they can be classified as earthfill dams, rockfill dams and concrete/masonry dams. 

The concrete/masonry dams can be classified further as gravity dams, buttress dams & arch dams based on their structural behavior and as conventional concrete dams & roller compacted concrete dams as per method of construction.

Conventional concrete dams are constructed by dividing the dam length into blocks of 20-25m long. Concrete placement is done by cableways, cranes, trestles etc. in lifts of 1.5-2m. The compaction of concrete is done by vibrators. Roller compacted concrete dams are constructed using same machinery/equipments as that used for embankment dams. Construction is done from abutment to abutment in lifts of 300-600 mm. Compaction of concrete is done with the help of vibratory rollers. 

Masonry dams were preferred in our country earlier as they were labour intensive, provided more employment opportunities, consumed less cement and did not involve any temperature control measures. However the quality of workmanship and workers are deteriorating now. There are problems of heavy seepage through many of our existing masonry dams. For seepage control, various remedial measures are being adopted these days, viz. guniting on upstream face, upstream concrete membrane, sandwich concrete membrane, prepacked masonry construction etc. Now-a-days, there is therefore a shift in favour of concrete dams. Further, the construction of concrete dams is faster vis-à-vis masonry dams. 

Gravity Dam

A concrete gravity dam is a solid concrete structure so designed and shaped that its weight is sufficient to ensure stability against the effects of all imposed forces. The complete design of a concrete gravity dam includes the determination of the most efficient and economical proportions for the water impounding structure and the determination of the most suitable appurtenant structures for the control and release of the impounded water consistent with the purpose and function of the project. 

General dimensions and definitions

Gravity dams may be straight or curved in plan depending upon the axis alignment. For uniformity, certain general dimensions and definitions have been established and are defined as below:

The structural height of a concrete gravity dam is defined as the difference in elevation between the top of the dam and the lowest point in the excavated foundation area. 

The hydraulic height is the difference in elevation between the lowest point of the original streambed at the axis of the dam and the maximum controllable water surface.  

The length of the dam is defined as the distance measured along the axis of the dam at the level of the top of the main body of the dam from abutment contact to abutment contact including the length of spillway if it lies wholly within the dam. However, the length of the abutment spillway located in any area especially excavated for the spillway is not included in the length of the dam.

The volume of a concrete dam includes the main body of the dam and all mass concrete appurtenances cot separated from the dam by construction or contraction joints. 

A plan is an orthographic projection on a horizontal plane, showing the main features of the dam and its appurtenant works with respect to the topography. A plan should be oriented so that the direction of stream flow is towards the top or towards the right of the drawing.

A profile is a developed elevation of the intersection of a dam with the original ground surface, rock surface or excavation surface along the axis of the dam, the upstream face, the downstream face or other designated location.

The axis of the dam is a vertical reference plane usually defined by the upstream edge of the top of the dam.

A section is a representation of a dam as it would appear if cut by a vertical plane taken normal to the axis and is usually oriented with the reservoir to the left.

Design Considerations

Local Conditions

Collection of data on local conditions will eventually relate to the design, specifications and construction stages of a dam.  Local conditions are not only needed to estimate construction costs, but may be of benefit when considering alternative designs and methods of construction.  Some of these local conditions will also be used to determine the extent of the project designs, including such items as access roads, bridges and construction camps.

Data required to be collected are:

i) Approximate distance from the nearest rail road shipping terminal to the structure site

ii) Local freight or, trucking facilities and rates

iii) Availability of housing and other facilities in the nearest towns

iv) Availability or, accessibility of public facilities or, utilities such as water supply, sewage disposal, electric power for construction purposes, telephone services etc.

v) Local labour pool and general occupational fields existing in the area

Maps and Photographs

Maps and photographs are of prime importance in the planning and design of a concrete dam and its appurtenant works.  From these data an evaluation of alternative layouts can be made preparatory to determining the final location of the dam, the type and location of its appurtenant works and the need for restoration and/or development of the area.

Data to be collected are:

i) A general map locating the area within the State, together with district and township lines.

ii) Map showing existing towns, highways, roads, railways and shipping points

iii) A vicinity map showing the following details:

a) The structure site and alternate sites

b) Public utilities

c) Stream gauging stations

d) Existing man-made works affected by the proposed development

e) Locations of potential construction access roads, sites for Government camp, permanent housing area and sites for Contractor’s camp and construction facilities

f) Sources of natural construction materials

iv) Site topography covering the area of dam, spillway, outlet works, diversion works, construction access and other facilities

Hydrologic Data 

In order to determine the potential of a site for storing water, generating power or, other beneficial use, a thorough study of hydrologic conditions is required.

The hydrologic data required include the following:

i) Stream flow records, including daily discharges, monthly volumes and momentary peaks

ii) Stream flow and reservoir yield

iii) Project water requirements, including allowances for irrigation and power, conveyance losses, reuse of return flows, dead storage requirements for power, recreation, fish, wildlife etc.

iv) Flood studies including inflow design floods and construction period floods

v) Sedimentation and water quality studies including sediment measurements, analysis of dissolved solids etc.

vi) Data on ground water tables in the vicinity of the reservoir and dam site.

vii) Water rights, including inter-state and international treaty effects.

Reservoir Capacity and Operation

Dam designs and reservoir operating criteria are related to reservoir capacity and anticipated reservoir operations.  The loads and loading combinations to be applied to the dam are derived from the several standard reservoir water surface elevations.  Reservoir capacity and reservoir operations are used to properly size spillway and outlet works.

Reservoir design data required for the design of dam and its appurtenant works are:

1) Area – Capacity curves and/or tables

2) Topographic map of reservoir area

3) Geological information pertinent to reservoir tightness

4) Reservoir storage allocations and corresponding elevation

5) Required outlet capacities of respective reservoir water surfaces and sill elevations etc.

6) Annual reservoir operation tables or charts

7) Method of reservoir operations for flood control, maximum permissible releases consistent with safe channel capacity

8) Anticipated wave action, wind velocity, fetch etc.

9) Anticipated occurrence and amount of ice, floating debris etc.

10) Physical, economic or, legal limitations to maximum reservoir water surface.

Climate Effects

Climate conditions at a site affect the design and construction of the dam.  Measures to be employed during construction to prevent cracking of concrete are related to ambient temperatures at site.

The data on climate conditions considered as part of design data are :

1) Records of mean monthly maximum, mean monthly minimum and mean monthly air temperatures at site

2) Daily maximum and minimum air temperatures

3) Daily maximum and minimum river water temperatures

4) Amount of annual variance in rainfall and snowfall

5) Wind velocities and prevailing direction

Construction Materials

Construction of a gravity dam requires availability of suitable aggregates in sufficient quantity.  Aggregates are usually processed from natural deposits of sand, gravel and cobbles or, may be crushed from suitable rock.

Data required on construction materials are:

1) Sources of aggregate

2) Water for construction purposes

3) Results of sampling, testing and analysis of construction materials

4) Information on potential sources of soils, sand and gravel to be used for backfill, road surface, protection of slope etc.

Site Selection

The two most important considerations in selecting a dam site are:

1) the site must be adequate to support the dam and appurtenant structures

2) the area upstream of site must be suitable for a reservoir

The following factors should be considered in selecting the best site out of several alternatives:

1) Topography : A narrow site to minimize amount of material in 

the dam, thus reducing its cost

2) Geology : Dam foundation should be relatively free of major 

faults and shears

3) Appurtenant : Selecting a site which will better accommodate the 

Structures appurtenant structures to reduce overall cost

4) Local conditions Sites requiring relocation of existing facilities like 

roads, railway, power lines, canals increase overall cost.

5) Access : Difficult access may require construction of 

expensive roads, thus increasing the cost.

Configuration of Dam

A gravity dam is a concrete structure designed so that its weight and thickness ensure stability against all the imposed forces.

Non-overflow section 

The downstream face is usually a uniform slope, which, if extended, would intersect the vertical upstream face at or near the maximum reservoir level. The upper portion of the dam must be thick enough to resist the shock of floating objects and to provide space for a roadway. The upstream face will normally be vertical. However, the thickness in the lower part may be increased by an upstream batter, if required. The base width (thickness) is an important factor in resisting the sliding and may dictate the d/s slope. 

Overflow section

Spillway may be located either in the abutment or in the dam. Section of spillway is similar to NOF section but modified at top to accommodate the crest and at the toe to accommodate the energy dissipater. The elevation of crest and its shape is determined by hydraulic requirements.

Foundation Investigation

The purpose of a foundation investigation is to provide data necessary to properly evaluate a foundation.  Basic data to be obtained during appraisal investigation, with refinement continuing until construction is complete are:

1) Dip, strike, thickness, composition and extent of faults and shears

2) Depth of overburden

3) Depth of weathering

4) Joint orientation and continuity

5) Tests of foundation rock viz.

Physical Properties Tests

- Compressive Strength

- Elastic modulus

- Poisson’s ratio

- Bulk specific gravity

- Porosity

- Absorption

Shear Tests

- Direct shear

- Triaxial Shear

- Sliding friction

Other Tests

- Solubility

- Petrographic Analysis

Construction Aspects

Construction aspects that should be considered in the design stage are:

- Adequacy of area for construction plant and equipment

- Permanent access roads to facilitate construction activities

- Length of construction seasons

Foundation Treatment For Dams

FOUNDATION TREATMENT FOR DAMS

 

1. INTRODUCTION

 

Dams are very important structures and have very stringent foundation requirements, viz., excellent bearing capacity, no differential settlement, almost impervious, perfect bond with dam material, no adverse effect due to earthquakes etc. Such a perfect foundation condition is seldom available, especially at locations where other conditions, i.e., hydrological, geological, seismo-tectonical, etc., are favourable for construction of a dam. Thus, we need to treat the existing foundation, so as to achieve geologic or geotechnical conditions that meet foundation performance requirements.

 

The preparation of the foundation and abutments for a dam is a most difficult and important phase of construction; the thoroughness with which it is done is reflected in the performance of the completed structure. It is often difficult or sometimes impossible to correct foundation and abutment deficiencies that show up after construction is well underway or completed. Hence, one need to take preventive actions by way of thorough geotechnical investigations and treat the foundation appropriately well in time, before it is too late and the safety & performance of the structure is compromised for years to come. 

 

About 59% of dam failures is due to foundation problems.

 ALSO READ: DESIGN OF CORE AND FILTER

 ALSO READ: DESIGN PRINCIPLES OF EARTH DAM

2. OBJECTIVES 

 

The primary objectives of foundation and abutment treatment are; 

 

1.     To provide adequate stability,

2.     To obtain positive control of under seepage, 

3.     To prepare surfaces to achieve satisfactory contact with overlying compacted fill, and, 

4.     To minimize differential settlements and thereby prevent cracking in the fill. 

 

These objectives can be achieved by defining foundation objectives, which serve as a descriptive tool to convey design intent information to construction engineering staff and the contractor. The general approaches available for defining foundation objectives are;

 

 Attain a specific geologic unit

 

An example of this approach is a narrative goal such as “extend the cutoff trench excavation completely through the alluvium and three feet into the underlying granitic rock.” This approach requires that sufficient exploration has been performed to identify a continuous geologic unit judged to possess adequate properties for the foundation.

 

 ALSO READ: DESIGN OF CORE AND FILTER

 ALSO READ: DESIGN PRINCIPLES OF EARTH DAM

 Excavate to a grade based on field testing results

 

Two examples of this approach are the quantitative statements “excavate the shell foundation to an elevation that encounters dense silty sand with an SPT N1(60) value of 30 blows per foot,” and “excavate the cutoff trench to an elevation that encounters crystalline rock with a low permeability of less than 10 lugeons.” This approach is typically used at well-explored sites, where a sufficient data set exists to identify the materials with stated properties across the site.

 

 Attain a specific rock quality

 

An example of this approach would be the narrative objective to “excavate the dam foundation to slightly weathered granitic rock.” Qualities of rock that are potentially significant to dam construction include degree of rock weathering and the density, orientation, aperture, and infilling of the discontinuities. This approach requires that sufficient exploration be performed to identify the consistent presence of the rock quality specified and reason to believe that rock with similar properties underlies the chosen surface.

 

 Achieve a surface that meets a construction control test

 

Two examples of this approach are “excavate to a surface with a relative compaction of 95% ASTM D-1557,” and “excavate to a surface with an in-place dry density of 120 pounds per cubic foot.” This approach is often used for poorly explored sites, where prejudgments cannot be made with confidence. Estimates of the excavation needed to achieve adequate foundation material can be poorly constrained, potentially increasing excavation costs. It requires an ability to physically test the foundation materials during construction, and a belief that materials with adequate properties will underlie the chosen surface.

 

 Excavate to a surface based on the ability of excavation equipment

 

An example of this approach is “excavate to blade refusal of a Caterpillar D-10N tractor dozer.” This approach usually requires a calibration test between the capability of the equipment, the character of the material on which it refuses, and adequacy of that material for foundation. This approach is not appropriate for stratified rock, where weaker materials may underlie a stronger layer.

 

 Excavate to a depth indicated by design analysis

 

This approach is often used when exploration indicates there is no expectation of material improvement within conventional excavation depth. The adequacy of the foundation is based on engineering analysis in conjunction with design solutions that mitigate the impact of the undesirable foundation materials.  Achieve a material judged adequate based on visual observation

 

An example of this approach is to “excavate to a depth directed by the engineer.” This approach requires the ability to make observations and judgments of strength and permeability during construction, and an expectation that adequate materials underlie the surface chosen. This approach is generally not used as the primary method of identifying adequate dam foundation materials, but should always be specified to confirm the adequacy of any surface indicated by other approaches, and to deal with unexpected foundation materials.

 ALSO READ: DESIGN OF CORE AND FILTER

 ALSO READ: DESIGN PRINCIPLES OF EARTH DAM

3. METHODS 

 

Foundation treatment methods depend on the actual site conditions, and as such, are as varying as the site conditions. The generalization brought out in this lecture is only intended to generalize approaches for treatment of foundation of dams for sake of easy comprehension.

 

Practicising engineers need to identify the potential problems and to suggest and adopt treatments, as appropriate on case to case basis.

 

 Clearing, Grubbing, Stripping, and Cleaning

 

Clearing consists of removal of all aboveground obstructions, including trees, vegetation, felled timber, brush, abandoned structures, local dams, bridges, and debris. Grubbing includes removal of all objectionable below ground obstructions or material including stumps, roots, logs, drain tiles, and buried structures or debris. Stripping consists of the removal of sod, topsoil, boulders, and organic or foreign materials. Shaping and cleaning consists of removing large loose rocks, overhangs, and projecting knobs by scaling, handpicking and wedging, and light blasting pressure washing followed by some form of “dental treatment” to fill all holes, cracks, joints, crevices, and depressions. These treatments are required to remove those materials having undesirable engineering qualities, such as, low shear strength, high compressibility, undesirable permeability, or other characteristics, which would interfere with compaction operations; and provide a surface favourable for a good bond with the overlying fill. Blasting should be avoided if possible; if unavoidable, explosive charges should be kept small as possible. The final preparation of almost all foundations should be by hand labor with adequate time given for inspections. The use of heavy or tracked vehicles on the final foundation should be avoided, especially if the rock is thinly bedded or badly jointed. 

 

 Seepage Control

 

Seepage control is by and large the most important foundation treatment measure. This is usually done by providing following;

 

 Cut-offs

 

Foundation cutoffs or core trenches serve as barriers to underseepage. The design of foundation cutoffs is based largely on borings made during field investigations for Detailed Project Report. Some common types of cut-offs are being described below;

 

(a) Compacted backfill trenches - Backfill compacted into a seepage cutoff trench is one of the most effective construction devices for blocking foundation seepage. Material and compaction requirements are the same as for the impervious section of the embankment. 

 

(b) Slurry trenches - The slurry trench method of constructing a seepage cutoff involves excavating a relatively narrow trench with near-vertical walls, keeping the trench filled with a bentonite slurry to support the walls and prevent inflow of water, and then backfilling with a plastic impervious mixture of well graded clayey gravel to protect against piping, to reduce seepage, and to minimize consolidation of the backfill material.

 

(c) Grout curtains - Grouting is the injection by pressure of grout (a mixture of water, cement, and other chemical compounds) into openings (voids, cracks, or joints) in a rock mass. The grout is designed to be injected as a fluid and to stiffen or solidify after injection. The rock foundation and abutments of most large dams require grouting to reduce seepage and to reduce hydrostatic uplift pressures in dam foundations. Grout curtains are frequently tied into the bottom of cutoff trenches which extend through soil overburden to the rock foundation. Grouting procedures must be tailored to the formation characteristics of the foundation being grouted, and close supervision and inspection are required to obtain satisfactory and economical results. 

 

(d) Upstream impervious blankets

 

A horizontal upstream impervious blanket controls underseepage by lengthening the path of underseepage. The effectiveness of the blanket depends on its length, thickness, continuity, and the permeability of the material/soil from which it is constructed.

 

 

 

Pressure relief wells, drainage galleries, toe-drains

 

Pressure relief wells, drainage galleries & tunnels, toe drains etc. are constructed to intercept seepage water, which might have passed through cutoffs, described above. Interception of underseepage water and relieving of excess uplift pressures prevents the transport of soil, which might occur in the formation of sand boils and also prevents heaving at the toe. 

 

(a) Pressure relief wells

 

Relief wells are installed along the downstream toe of an embankment to intercept underseepage water and relieve excess uplift pressures that would otherwise develop at the toe of an embankment. 

 

(b) Drainage galleries and tunnels

 

To facilitate foundation and abutment grouting and interception of seepage water, drainage galleries and tunnels are provided in high dams. 

 

(c) Toe drains

 

Toe drains collect and facilitate removal of seepage water at the downstream toe of the dam to prevent formation of soft boggy areas and/or boils. Toe drains are generally connected to the horizontal drainage blanket and sometimes to the relief well system to collect and remove seepage water in thin pervious strata in the upper foundation that the deeper relief wells cannot drain.  Treatment of Unfavourable Conditions

 

Unexpected unfavorable conditions are frequently discovered during early construction, and may range from undesirable deposits of material not detected in exploratory drilling to adverse seepage conditions that were impossible to predict. Some common undesirable conditions are discussed below;

 ALSO READ: DESIGN OF CORE AND FILTER

 ALSO READ: DESIGN PRINCIPLES OF EARTH DAM

3.3.1. Unfavourable soil conditions

 

(a) Highly compressible and low strength soils - Organic soils exhibit high compressibility and low shear strength and are generally recognized by their dark color, the presence of organic particles, and often a distinctive “organic” odor. Inorganic clays with high water content also exhibit high compressibility and low shear strength. If an embankment is constructed on a deposit of either highly organic soil or highly compressible inorganic soil, excessive differential settlement could cause cracking of the embankment, or shear failure might occur; if significant deposits of either of these materials are discovered during early construction, their extent should be established and, if it is feasible, they should be removed and replaced with acceptable compacted backfill. 

 

(b) Clay shales – Clay shales are among the most troublesome and unpredictable soils. They are often termed “compaction” or “soil-like” shales, if they have been highly over consolidated by great thicknesses of overlying sediment and have no appreciable cementation. Clay shales tend to slake rapidly, when subjected to cycles of wetting and drying; some exhibit very high dry strength, but upon wetting swell and slake profusely, losing strength rapidly. They swell or expand considerably when unloaded by excavating overlying material due to release of stored strain energy. Therefore, excavating in clay shales should be completed and backfilled without delay. The last foot or so of excavation into slaking clay shale should be deferred, until just prior to backfill operations in order to minimize the time of exposure of the final clay shale surface. 

 

(c) Collapsible soils - “Collapsible” soils are generally soils of low density and plasticity, which are susceptible to large decreases in bulk volume when they are exposed to water. Collapsible soils are characterized by bulky grains (in the silt-to-fine sand grain size) along with some clay. Collapse results from softening of clay binder between larger particles or the loss of particle-to-particle

 

 

cementation due to wetting. Volume change from collapse occurs rapidly (relative to consolidation) and can be very significant, especially if the soil is under high stress.

 

(d) Loose granular soils - Loose, water-saturated sands and silts of low plasticity may have adequate shear strength under static loading conditions; however, if such materials are subjected to vibratory loading, they may lose strength to the point, where they flow like a fluid. The process in which susceptible soils become unstable and flow when shocked by vibratory loading is called liquefaction, and it can be produced by vibration from blasting operations, earthquakes, or reciprocating machinery. In very loose and unstable deposits, liquefaction can occur as the result of disturbances so small that they are unidentifiable. Such loose silt and sand deposits may be compacted by blasting (generally not effective in densifying loose granular deposits, because the vibratory energy produced is of such high frequency), vibroflotation, and driving compaction piles; however, the effectiveness of these procedures for deposit densification is not predictable.  (e) Steep abutment slopes - Steep abutment slopes of earth tend to increase the possibility of transverse cracks developing in the embankment after construction. During construction, they may become unstable and endanger construction personnel. Slides can occur in clays, sands, and gravel, particularly in slopes subjected to seepage. Slides may damage completed works and require costly repairs. In many cases, it may be necessary to bench the slopes to provide safety against sloughing material and sliding. 

 

(f) Old river channels - Old abandoned river channels filled with pervious or impervious materials are often encountered unexpectedly during construction. As mentioned earlier, the extent of these deposits is often difficult to establish accurately during the exploratory stages, and in some cases an entire deposit may be missed. Where the existence of such deposits has been revealed, additional exploration by borings, test pits, etc., to establish their extent may be necessary. Old river channels beneath a dam foundation, filled with course-grained pervious material, would constitute a dangerous open path of seepage. Channel fillings of soft fine-grained materials can cause differential settlements and cracking of the embankment, if not removed and replaced with properly compacted material. 

 

3.3.2. Unfavourable rock conditions

 

(a) Weathered rock - Weathered rock may have undesirable characteristics, such as, high compressibility, low strength, and high permeability. Removal of weathered rock is generally required for embankments founded on rock to obtain impervious contact beneath the core and to eliminate the possibility of differential settlements and low shear strengths beneath the core and other zones. The weathering of rock is a transitional process; a sharp line of demarcation does not exist between weathered and unweathered zones.

 

In fact there are nine categories of weathering, as under;

 

(i)    Fresh (Intact rock)       – W1; 

(ii)   Slightly weathered to Fresh      – W2; 

(iii)  Slightly weathered (shallow oxidation or discolouration)   – W3; (iv)  Moderately to Slightly weathered     – W4;

(v)   Moderately weathered (significant oxidation or discolouration,         body of rock slightly weakened, open joints)    – W5;

(vi)  Intensely to Moderately weathered     – W6; (vii) Intensely weathered (thoroughly oxidised or discoloured,         body of rock significantly weakened)     – W7;  

(viii)Very Intensely weathered (joints fully separated)   – W8;

(ix) Decomposed (more like soil)      – W9.,  The degree of weathering usually decreases with depth, thus, it may be necessary to excavate deeper in some areas than in others to remove weathered rock beyond category W2. 

 

(b) Open joints and fractures -  All open joints, cracks, fissures, and fractures in the foundation rock surface must be filled to prevent erosion or scour of embankment material at the rock contact. A sand-cement mortar is generally used to fill these openings. 

 

 

(c) Cavities and solution features - Cavities, potholes, and other voids caused by solution of the rock are dangerous, and field personnel should always be on the lookout for such conditions during foundation preparation. More care should be taken, where a dam is being built on rock susceptible to solution, such as limestone or gypsum. Potholes and cavities exposed or “day lighted” on the foundation surface are usually remedied by dental treatment. Concrete should be thoroughly vibrated & rodded into the voids and its upper surface brought up to the general level of the surrounding rock. Dental treatment serves to smooth up the foundation to reduce compaction difficulties as well as provide a non-erodable impervious seal, a measure of protection against scour of the embankment fill along the rock contact. 

 

(d) Overhangs and surface depressions - Overhangs and other irregularities in the rock surface of an abutment or foundation must be corrected. Overhangs should be removed by drilling and blasting, preferably with pre-splitting, so as not to disturb the adjacent sound rock. Concrete dental treatment can be used to fill depressions created by blasting and to remedy some types of overhangs. Tamping of soil under overhangs instead of removal or dental treatment must not be permitted. If the rock is very irregular, it may be more economical to cover the entire area with a concrete slab. It should be noted that a gently undulating rock surface is desirable, and only when surface depressions interfere with compaction of the first lift, should concrete backfilling be required.

 

(e) Springs - Springs, often encountered in rock foundations and abutments, are simply groundwater sources seeping to the ground surface driven by artesian pressure. Attempts to place impervious fill over springs issuing from joints or rock fractures will result in extremely wet soil in the vicinity of the spring, which is impossible to properly compact. Depending on the flow rate and pressure driving the spring, seepage will continue through the wet soil, creating an uncompacted weak zone of increasing size, if fill placement is continued without properly removing this source of water. The zone created around an improperly controlled spring is a very dangerous situation, which will cause problems both during construction and over the life of the embankment. Where the water is under a low head and has a single point of issue, a standpipe can usually be installed. A corrugated metal pipe of a diameter depending upon the size of the spring is placed over the spring area, and a damp mixture of quicksetting cement, sand, and gravel is packed around the standpipe base. Earth is then compacted around the outside of the pipe at the base. The water is kept pumped down within the standpipe until an impervious seal is obtained and enough pipe sections have been added to retain the head of water in the pipe. The pipe is then filled with vibrated concrete or grout, and construction of the fill continued upwards and across the top of the plugged pipe by conventional methods. The area is then examined for evidence of new springs, which often appear after an old spring is plugged. This procedure can also be used for springs on the abutment when the fill reaches the same elevation as the spring. While filling operations are progressing below the spring, a small pipe can be grouted into the source of seepage and discharged away from the fill as a temporary measure. 

 

 Dewatering and Drainage of Excavated Areas

 

Inadequate control of groundwater seepage and surface drainage during construction can cause major problems in maintaining excavated slopes and foundation surfaces and in compacting fill on the foundation and adjacent to abutment slopes. Dewatering systems should be adequate enough to control seepage and hydrostatic uplift in excavations, and for collection and disposal of surface drainage and seepage into excavations.  

 

 Dewatering - Potential troubles can often be detected in early stages by visual observation of increased seepage flow,  piping of material from the foundation of slopes, development of soft wet areas, uplift of excavated surfaces, lateral movement of slopes, or failure of piezometer water levels to drop sufficiently as pumping is continued. Water pumped from dewatering systems must be observed daily at the discharge outlet; if the discharge water is muddy or contains fine sand, fines are being pumped from the foundation. This is important as the

 

 

pumping of fines from the foundation can cause internal erosion channels or piping to develop in the embankment structure; if this happens, it is crucial that corrective measures be taken.

 

Failure of the dewatering system can result in extremely serious problems, often requiring extensive and expensive remedial work. In excavations bottoming in impervious material, unchecked artesian pressure in underlying pervious strata can cause heaving of the excavation bottom. If the impervious stratum ruptures under these pressures, boils (violent emission of soil and water) will develop, causing the loss of the underlying foundation material and thereby endangering the entire structure. Similar boils could develop on the bottom of an excavation from excessive artesian pressures in the underlying strata. Failure of excavation slopes may also occur because of excessive artesian pressures. In order to prevent failure of the dewatering system, all power sources should have standby gas or diesel-powered pumping or generating equipment, and standby pumps should be available.

 

 Sumps and ditches - When an excavation, such as a cutoff trench, is extended to rock or to an impervious stratum, there will usually be some water seeping into the excavation and/or “wet spots” in the bottom of the excavation, even with deep wells or well point systems in operation. Water seeping into the excavation from the upstream and downstream slopes of a long cutoff trench can usually be captured by excavating narrow longitudinal ditches or drainage trenches at the intersection of the slopes and the bottom of the excavation, or by forming such trenches with sandbags, with sumps located as necessary for pumping the water out. If the bottom of the excavation will still not dry out, smaller ditches can be cut through the problem areas and sloped to drain to the side trenches.

 

 Surface erosion - Surface erosion may present problems on slopes cut in silts, fine sands, and lean clays. Eroded material will wash down and fill in the excavation below the slope. The slope itself will be left deeply scoured and rutted, making it necessary for costly smoothing operations to be performed before the fill can be placed against it. The best way to combat surface erosion of temporary excavation slopes is to backfill as soon as possible, thus cutting down on exposure time. This often cannot be done, however, and it becomes necessary to take other measures. Grass cover on the slopes is a good means of preventing surface erosion, if it can be readily established and if the slopes are to remain open for a season or two. Other slope protection measures such as rip-rap or asphaltic treatment are rarely justified for construction slopes. Thus, it is necessary to keep as much water off the slope as possible. Most slopes can withstand rain falling directly on them with only minor sloughing. Perimeter ditches and/or dikes at the top of the slope are needed to carry other surface waters away from the excavation, if surface waters outside the excavation would otherwise run into it. Ditches may be needed at several elevations along the excavation slopes to catch surface waters.

 

 Other seepage control measures - Other means of stabilizing excavation slopes and preventing seepage from entering an excavation (such as electro-osmosis, freezing, sheet-piling, and grouting) have been used for structure excavations. These methods are not economically feasible for extensive foundation excavations for dams, but might be used in structures, where conventional dewatering methods are not suitable for various reasons.

 

 ALSO READ: DESIGN OF CORE AND FILTER

 ALSO READ: DESIGN PRINCIPLES OF EARTH DAM